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Re: Plywood rigid diaphragms
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- Subject: Re: Plywood rigid diaphragms
- From: Lynn <lhoward(--nospam--at)silcom.com>
- Date: Sat, 29 Aug 1998 08:08:52 -0700
Michael- Thanks for a good run down on this subject. I appreciate it. What about plywood shear walls that do not have any holdowns? (there is no calculated uplift.) And what if those wall are in line with a wall that does have a holdown? How about Simpson PHDA holdowns that supposedly limit holdown deflection? How about strap type holdowns? Will a strap with 500 lb. uplift have the same deflection as an HD20A with 15,000 lb. uplift? Do you need to calculate the deflection of each holdown type based on your guess of the loads, run the analysis, and then re-run it? And when the Architect moves a door or window at the last minute do you tell him it will cost $2,000 to re-design the walls? You won't necessarily be calculating the deflection of a single wall, but the deflection of a "wall line". But another problem is that these plywood walls do not distribute loads like concrete walls do. Due to slippage of nails and other factors, the forces tend to equalize themselves along the run of a wall. The deflection at the top of the wall at one point may not be the same as another. The horizontal diaphragm must also be "flexible" enough to accommodate this re-distribution of stresses or the whole theory goes out the window. In my opinion, this falls in the category of trying to fix something that is not broke. From what I have seen and read about the performance of wood framed structures during recent earthquakes, wood framed horizontal and vertical diaphragms did very well with the exception of tall slender walls. I tend to lean toward the "where are the bodies" camp when making significant code changes. You are really going to open a pandora's box if we try and go down this road too far. Near as I can tell, most buildings perform fine when horizontal and vertical diaphragms are considered flexible. Lynn Mlcse(--nospam--at)aol.com wrote: > Los Angeles has adopted a new voluntary ordinance called Division 93: > Voluntary Earthquake Hazard Reduction In Existing Wood Frame Residential > Buildings with Soft, Weak or Open Front Walls. In this ordinance, you are > required to calculate shear wall deflections to show that the deflection is > less than .005H. The shear wall deflection equation is from the UBC Volume > 3. > > Using this equation, the biggest part of the calculated shear wall deflection > is due to holdown slip (minimum 1/16" oversized hole in post for holdown > bolts, deflection of manufactured holdown base, shrinkage of lumber, crushing > of wood sill under vertical loads, etc). Because of the holdown slip, the > plywood diaphragm will be rigid as compared to the the plywood shear walls. > The rigidity of the plywood diaphragm is going to primarily be based upon how > rigid the walls are. In commercial buildings with masonry or concrete shear > walls, or steel braced frames, the deflection of the walls or braced frames > will be small compared to the plywood diaphragm, therefore the diaphragm will > behave flexibly. > > When you have wood shear walls, typically the architectural layout does not > allow for many walls because of all the openings for doors and windows. In > residential construction (single family homes, apartments, small commercial, > etc), parallel shear walls may not be very far apart, say 30 or 40 feet, and > then the diaphragm length is 40 to 80 feet in the other direction. When you > compare the deflection of the most likely few shear walls along each edge of > the floor diaphragm with the deflection of the floor plywood diaphragm you > will probably see that the floor should be considered rigid. > > Example: For an 8 foot tall plywood shear wall (platform framing), the > allowable deflection is .005H = 8 x 12" x .005 = 0.48 inches under design > loads. For an 8 foot tall masonry or concrete shear wall, wouldn't you expect > the deflection to be much less, say 0.05 inches. With this type of shear wall > deflection, it may be easier to visualize how a plywood diaphragm is > considered to be rigid compared to the more flexible wood shear walls. > > This becomes a nightmare when you try to distribute the load by rigidity > though, since you do not know the rigidity of the plywood shear walls, > primarily because of the holdown slip and certain walls have more gravity > loads to resist overturning as opposed to other walls. This then becomes an > iterative solution, since as you change the loading to a wall you change the > holdown size and therefore the holdown slip and deflection. There are also > those codes which do not allow you to transfer forces by rotation when using a > wood diaphragm, sort of a catch 22, the plywood floor is rigid, but you can't > distribute the loads by rigidity. > > The holdown movement component for deflection (vertical displacement x (wall > height / wall length)) is the main component of the total shear wall > deflection. On a 4x8 shear wall panel, the horizontal deflection due to just > the 1/16" oversized hole in the post is 0.125" (1/16 x (8/4) = 0.125") which > is 26% of the allowable for an 8 foot tall wall (0.125"/0.48" = 26%). When > you add in wood shrinkage and holdown deflection, the deflection increases > furthur and you have not yet included the deflection from the rest of the > shear wall deflection equation ( wall bending, wall shear, nail slip). The > situation becomes even worse when you calculate the deflection for the second > story shear walls, because of more deflection and slip of the holdown > assembly, now sitting on the wood floor instead of the concrete foundation. > > Basically it seems the traditional holdowns do not work on short walls when > you try to limit the deflection to 0.005H, the wall has already exceed the > allowable deflection before the holdown sees any load. In order to meet the > deflection criteria of 0.005H, the shear walls will have to probably be > longer, say 1:1 or 1:2 even for multistory buildings which won't go over very > well with architects. The proprietary wood shear walls (Simpson Strong-Wall, > Shear Transfer Systems STS panel) appear to work well for one story > applications where wall aspect ratios are 2:1 or less since there design > values are based upon tests. But, I personally would be cautious about using > them for a second floor shear wall, unless you reduce the allowable design > loads to account for the change in wood floor stiffness as compared to the > wall sitting on a concrete foundation. > > I think the concerns are that if you design strictly by tributary area, you > may end up underdesigning certain walls. In a simple box shaped building with > perimeter shear walls only, the short wall on one side is designed for 50 > percent of the tributary area, the long wall on the other side is designed for > 50% of the tributary area. The short wall deflects (never sees the full 50%) > and the long wall now sees 70-100% of the load when it was actually just > designed for 50% of the load. Now the question is how to come up with a > reasonable solution, which can be implemented within the design fee structure > that as engineers we typically get for designing wood structures. > > Any suggestions? > > By the way, SEAOSC is planning a seminar on this topic for November. > > Michael Cochran
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