Return to index: [Subject] [Thread] [Date] [Author]

RE: AISC Seismic Provisions

[Subject Prev][Subject Next][Thread Prev][Thread Next]
The requirements for CBF's are intended to eliminate local buckling of the
member since this may lead to fracture and brittle failure, which is
undesirable. That is why you end up with a member that has smaller b/t ratio
compared to an ordinary braced frame.

In regard to your question about the criteria for connection design; the
intent here is to make the yielding start in the brace (overall buckling)
rather than the connection. Hence the connection strength should exceed the
ultimate tensile capacity of the brace. Or, If the system, meaning your
diaphragm, collectors, shear transfer, etc. is capable of transferring a
smaller force to the brace, you only need to design the connection for that
force. This is not the force that you get from your analysis. You need to
look at each of these elements that deliver the force to the brace and see
which one will govern (the smallest force) and compare that to the tensile
capacity of the brace.

Many designers do not bother looking at the system forces and just design
for the tensile capacity of the brace and that is OK.

Ben Yousefi, S.E.


	-----Original Message-----
	From:	Connor, John A NWK [SMTP:John.A.Connor(--nospam--at)nwk02.usace.army.mil]
	Sent:	Friday, April 09, 1999 1:30 PM
	To:	seaint(--nospam--at)seaint.org
	Subject:	AISC Seismic Provisions

	I would appreciate any comments and clarifications on the following
problem.

	Design code:  1997 NEHRP (seismic provisions), 1997 AISC (steel
building
	seismic provisions), and ASCE 7-95 (wind and dead/live loadings)
	Project Location:  Kansas
	Building type:  Aircraft hangar

	I am designing a steel braced frame for an aircraft hangar using the
codes
	listed above.  My problem is determining the required capacity for
the
	bracing connections.  

	NEHRP directs me that my seismic requirements shall be in accordance
with
	AISC's Seismic Provisions for Structural Steel Buildings.  Based
upon the
	significance of this building, I have decided to design the steel
framing to
	meet the requirements of AISC's chapter 13, "Special Concentrically
Braced
	Frames".

	The braced bays of this building consists of x-braces.  The bracing
members
	are approximately 50' in length.  Assuming I use a steel tube shape,
chapter
	13's requirements for slenderness and also limiting width-thickness
ratios
	leads me to select a TS 10x6x5/8.

	I used this size as my preliminary member size in my STAAD model.
After
	running all of the different load cases, the output for this member
is about
	60 kips for both compression and tension.  The "controlling" load
case for
	this output was from the wind loads.

	Chapter 13 says that the required strength of my bracing connection
shall be
	the LESSER of the following:

	a)  "The nominal axial tensile strength of the bracing member,
determined as
	RyFyAg."
	For this tube: (1.1)*46*16.4=830 kips.

	b) "The maximum force, indicated by analysis, that can be
transferred to the
	brace by the system."
	I interpret this as my STAAD output which equals 60 kips.

	According to chapter 13, the "lesser" would be a connection designed
for 60
	kips.

	Did I interpret the criteria correctly?  Did I miss the "fine print"
	somewhere?

	What I don't understand is, why do I have to have such a "huge"
bracing
	member, when I only have 60 kips going through the member?  In the
event of
	a seismic event, my connection would fail before the member would
"buckle"
	or "snap".  My seismic event forces are less than my wind forces.
It
	doesn't seem right requiring a large brace and only have a 3-bolt
	connection.

	Also, when would condition "a" control over condition "b"?  For
example, if
	the analysis says my maximum tension is 500 kips, I would have to
select a
	member for (0.9)*Fy*Ag.  0.9<1.1 so I would never have this as a
controlling
	case.  Does the code have a mistake in wording, or is it correct?

	Any information or comments would be appreciated.

	John Connor, EIT
	Kansas City, MO