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RE: AISC Seismic Provisions for SCBFs

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You must keep in mind that the performance in a seismic event is different
than the performance in wind resistance. For a wind event, the structural
response is linear (elastic).  For seismic resistance, if you use the
standard response coefficients, you are assuming that the structural
response will be nonlinear.  The nonlinear area is where the seismic energy
is dissipated.  The only way to assure nonlinear performance is to employ
the special seismic provisions for the connections, braces, etc.

Intermediate seismic zones and high wind & high seismic zones are tricky in
this area in that a member may have more force in it due to wind, but you
may have to design the connections due to seismic.  In essence you design
the structure twice.  

You could use a low R or Rw value (like unreinforced masonry) and get to an
elastic seismic performance, but then you probably would see the pseudo
seismic member loads exceed the wind loads (even in intermediate and low
seismic zones).

As a former Midwest (Kansas City) designer, I empathize with your problem.
Seismic design is not taught in the area universities, and the concepts can
be a bit foreign.

You could wait to design in Kansas City until the IBC is adopted.  Kansas
City will have its seismicty reduced in the IBC.

Harold Sprague
The Neenan Company

-----Original Message-----
From: Connor, John A NWK [mailto:John.A.Connor(--nospam--at)]
Sent: Thursday, April 15, 1999 1:35 PM
To: 'seaint(--nospam--at)'
Subject: RE: AISC Seismic Provisions for SCBFs

I appreciate all of the responses to this subject.  All of the responses
helped to clarify some the terms and wording used in the AISC codes.  I
understand the concept of having the brace yield before the connection,
however, I don't understand why I have to meet certain criteria when I
expect an elastic response versus a deformation response.

For example, for most of the Midwest or areas of low seismic forces, wind
forces would generally "control" over the seismic forces.  Since my wind
forces control, my members are designed for elastic responses.  These
members wouldn't be able to reach deformation levels since my seismic forces
are so low.  Since I expect the members to remain elastic, I believe these
members would be considered "force-controlled" members rather than
"deformation-controlled" members.  

I originally thought my problem was determining the strength of the
connection ("brace strength" vs. "transferred by the system").  Now that I
know what "transferred by the system" really means, I find a problem
accepting the criteria for the limiting KL/R ratios.   The member I'm using
for the brace was not selected based on its capacity.  The member was
selected based on the limiting KL/R ratios.  If the KL/R ratios wasn't so
limiting, I wouldn't need a TS10x6.  A smaller TS size would be satisfactory
with regard to capacity.

I've read through the commentaries and the responses to this list concerning
the concept behind why the ratios are limiting, and I agree with the concept
behind it.  However, since I'm expecting elastic responses from my system,
why do I have to meet the KL/R criteria, when the criteria applies to
inelastic responses?

Is there some "fine print" somewhere that says the criteria is only
applicable to deformation/inelastic responses?  At the beginning of AISC's
seismic provisions, it says that the criteria applies to seismic design
categories D or greater.  My building happens to be category "D" so the AISC
criteria apparently applies.

John Connor, EIT
Kansas City, MO