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RE: Seismology Opinion - Cantilever Columns

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Bill,
I will be there. Bill Warren called me about this two weeks ago and I sent
him a brief outline of my portion of the seminar the following Sunday. I
sent a copy to all the people on his distribution list so I thought you
knew.
Thanks
Dennis

-----Original Message-----
From: William Nelson [mailto:BNelsonSE(--nospam--at)email.msn.com]
Sent: Tuesday, February 08, 2000 1:21 AM
To: seaint(--nospam--at)seaint.org
Subject: Re: Seismology Opinion - Cantilever Columns


Dennis,

I wrote the Opinion and will be more than happy to discuss the decision
process that was involved.  Our meeting for the March 25 seminar will be
Green River golf course 3:00 PM, 2/8/00.  I sent you an e-mail about it last
week but have not heard back yet... can you be there?

Bill Nelson

BNelsonse(--nospam--at)msn.com
    or
Bill(--nospam--at)jnase.com


----- Original Message -----
From: SEConsultant <seconsultant(--nospam--at)earthlink.net>
To: <seaint(--nospam--at)seaint.org>
Sent: Saturday, February 05, 2000 1:27 PM
Subject: Seismology Opinion - Cantilever Columns


> I just read the Seismology Opinion for the use of Cantilever Columns. The
> provisions are not too bad until you get to issue number 5 which state:
> "The total shear in all columns combined shall be less than 15% of the
story
> shear based on tributary area."
>
> I interpreted the document to read:
>
> You may design by factoring up the loads to the embedded columns only IF
and
> only if these provisions are met - otherwise, you must factor up the
entire
> structure in the direction of force to match the lower R.
>
> Let's look at a couple of conditions to see how restrictive this actually
> is:
>
> 1. It eliminates most conditions where embedded columns are used between
> lines of adjacent shear (assumed 50% of the tributary force) or at either
> end of a series of diaphragms (assumed to be one) resisted by three lines
of
> shear (approximately 25% of the load at the ends) or any condition that
does
> not result in a very small tributary width totaling no more than 15% of
the
> story shear.
>
> This is a very restrictive clause which will require most structures to be
> penalized by forming them into compliance with a total lowered R in the
> direction of applied load.
>
> In my opinion, limiting the rule by a percentage of the base shear is too
> generic a method of determining appropriate application because it tends
to
> eliminate most conditions other than the edge of a patio structure. If
this
> was the intention, why not just restrict it this way rather than trying to
> do it mathematically?
>
> What appear rational, is the restriction of story drift to 0.005H or "the
> approximate deflection of the adjacent shear walls in the same orthogonal
> directing". I agree with this - it is rational and insures a uniform
> stiffness - it makes sense. 15% of total base shear makes no sense.
>
> The opinion also limits the column axial stress ratio based on a K=2.1
> (which is typical of a column allowed to rotate and translate at the top
> while fixed at the bottom) to be less than 10% of the allowable axial
stress
> (fa/Fa). This does not makes sense, simply because the steel section
> required to resist the story drift restriction will make this a moot
point.
> In most cases, in lightweight wood construction, the axial capacity of the
> column will be many time greater than the actual load.
>
> Let's look at a simple 10' Pipe column which is designed for a one kip
> lateral load. The factored load will be 2.5 kips (neglect axial load for
the
> moment). The minimum size of a pipe column to resist the load will be P8xs
> (8" diameter extra strong schedule 40 pipe column).  This is required to
> keep the story drift below 0.6" at 0.005H (a P8std will deflect about
0.685"
> exceeding 0.005H). The allowable Axial Stress Fa is around 22.7ksi for
> DL+Short Term loading. This is where the axial ratio gets kind of
ridiculous
> because the columns would be allowed up to approximately 24.3 kips just to
> meet 10% of their capacity.
>
> I'm trying to understand how these restrictions were arrived at - maybe
the
> Seismology committee can provide notes as to what rationale led to these
> restrictions so, as designers, we can understand what the group is
> attempting to protect. This example uses a relatively small lateral load
of
> only 1000 pounds. The load to a front of a 22' deep garage caused by wind
> loads is closer to 2000 pounds of shear or 5 kips applied to two columns
> (factored 2.5 times).
>
> Now here is the next part that confuses me:
>
> In my mind the stiffness of the resisting element has less to do with it's
> material than the ability to calculate the actual deflection under a given
> load. Therefore, why does it matter whether we are dealing with an
embedded
> steel column, a plywood shearwall or a masonry wall. As long as we can
> calculate the deflection on each element we can compare performance under
> load. The stiffness or rigidity becomes a comparison between calculated
> deflections in each line of shear.
>
> If the goal is to balance deflection - essentially equalizing stiffness
> between lines of shear - why would we want to factor uniformly through the
> structure. I would think that we would want only to factor up the loads to
> elements that are normally more flexible in order to make them converge on
> the same stiffness as the more rigid elements.
>
> I would like to read the rationale on how this Opinion was formulated. Any
> suggestions on how this can be accomplished?
>
> Regards,
> Dennis S. Wish, PE
> Structural Engineering Consultant
> (208) 361-5447 E-Fax
>
>
>