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Near future of Steel Moment Frames

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Our office does quite a bit of design using moment frame construction
(chiefly SMRF's).  As such, we have closely followed the research and
documents on the subject.  Nearly all of our design incorporates the use of
W24 or larger columns, so we have regularly asked about the progress of
testing of these larger sections.  We have received little response from
those "in the know".  One friend of mine stated that during a test on a W24
column they encountered instability, which they attributed to the kinking in
the flanges due to panel zone deformation.  The column was reinforced per
the current "minimum strength" provisions shown in the UBC and FEMA
guidelines.

Stephen P. Schneider wrote an article for the Journal of Structural
Engineering which was printed January, 1998, which showed a methodology for
estimating the increased seismic drift due to panel zone deformation in a
structure.  This has been adapted and included in our office's moment
connection software.  On the first project
that we used this on, we calculated a 50% increase in the drift due to panel
zone deformation.  (Note that this means that about 33% of the total drift
is coming from the panel zone).  We were obviously alarmed and sent all our
information to Dr. Schneider for validation.  He confirmed our findings
were consistent with his experimental findings (that the panel zone
deformations constitute almost 1/3 of the total joint deformations.)  With
this knowledge, we have instituted a practice of checking both strength and
stiffness of panel zones, which is done in nearly every other area of
engineering (ie: beams, walls, frames).

A copy of the SAC 100% draft of FEMA 350 obtained by our office (not off
this list server, and perhaps not the latest draft) indicated that W24 and
larger columns were not pre-qualified for all
connection types that rely on panel zone participation to achieve required
rotational capacity.  (I guess SidePlate are the only ones not sweating
now.)  We are fairly alarmed, because the cost of designing frames with W14
columns is ominous.  I also don't want to be the one to tell an owner that
we will need to do full scale testing to prove that a connection is viable.
Additionally, if we are adhering to this philosophy, and the other local
engineers are not, we will be called to the table for conservatism (Why is
your building 20 psf, when the last one we did was only 15 psf?)  I
understand that we would not be forced to comply with these recommendations
unless the building official requires it;   However, we also cannot design
below our own level of confidence and hope to hide behind the building code.

I am beginning the engineering on a simple 6 story SMRF office building, and
would like the input of those on the server one this issue before I have an
uncomfortable meeting with the owner.  I would be particularly interested in
any ideas from those involved in the preparation of the FEMA 350 document
from SAC, AISC, or any other individual all the way down to those in the lab
running the tests.  Just how safe and effective are W24 or larger columns in
moment frames?  Is there enough test data to prove that they have problems?
What testing is currently being done to prove whether or not that they are
viable?

........Kimberley Robinson........
.......Dunn Associates, Inc........
Consulting Structural Engineers
801.575.8877   801.575.8875




........Kimberley Robinson........
.......Dunn Associates, Inc........
Consulting Structural Engineers
801.575.8877   801.575.8875



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