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RE: 1997 UBC 2320.2 & 1605.2

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Steve,
Assuming that the condition occurs in one location (in one line of
shear) I would assume the R for the home as 5.5 and adjust the one line
of shear in the following manner:
1. At the upper floor where you are using (I assume) proprietary braced
frames - I would adjust the shear in this line for an R of 4.5 - i.e.;
take the tributary shear at this line and multiply it by 5.5/4.5 to
adjust its stiffness.
2. The uplift and compression generated from the braced panels need to
be designed into the GLB below. The problem is that the flexibility of
the GLB may contribute to greater deflection in the braced panels above.
As I recall, the Seismic Design Manual Volume II had a wood problem
where a shearwall was placed on top of a wooden beam. You might want to
refer to this and find out if there were any corrections to the method
posted after publication.
3. I would then take the accumulated shear at the second floor diaphragm
(roof and 2nd floor shear)from the original analysis which used R of 5.5
and adjust this for an R of 2.2 to design the cantilevered column.

This is how I would design each step from roof down to foundation,
however, while I would do this to satisfy code (or SEAOC's Blue Book
opinion on cantilevered columns) I don't really understand their intent
in penalizing a structure because it was not common practice in the past
to design plywood shearwalls, or embedded columns to deflection limits.

Just for discussion purposes, it does not make much sense to me to
adjust the R value for each member design. I understand when making a
connection of steel to steel that increasing the Rw value from the old
UBC (94 version) assured a sufficiently larger factor of safety in the
design of the connection. However, with embedded or cantilevered
columns, the stiffness issue is moot if you are designing the member for
deflection rather than allowable bending. What I mean is that while you
may still be within the allowable stress range for bending, the column
can still deflect more than the code limit (assuming 0.0025H). This was
the problem that occurred when engineers designed high load plywood
shearwalls by assuming that they only needed to comply with a 3.5:1
aspect ratio (H/b). When these walls were studied after failures, it
became apparent that the allowable wall deflection occurred at a much
lower capacity than the wall was sheathed for.

So why not just require the design of the plywood wall, braced frame or
cantilevered column within an allowable deflection limit based on the
initial demand calculated to the entire building (say R of 5.5)? What
benefit does it serve to design the columns to an R of 2.2 when it only
makes them much stiffer than they need to be?

Does this make much sense? It is understandable to me when designing a
connection - but not when designing a shear resisting element to a set
deflection limit.

Dennis S. Wish, PE
California Professional Engineer
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-----Original Message-----
From: Steve Hiner [mailto:shiner(--nospam--at)folsom.ca.us]
Sent: Friday, February 08, 2002 12:04 PM
To: 'seaint(--nospam--at)seaint.org'
Subject: 1997 UBC 2320.2 & 1605.2

Section 2320.2/Design of Portions - essentially states that
"nonconventional" structural elements are to be designed in accordance
with
Section 1605.2

Section 1605.2 requires a "rational analysis in accordance with well
established principles of mechanics ... etc."

Here's the problem:  two story wood frame residence with braced wall
panels
in the 2nd story that are not continuous through the 1st story to the
foundation (i.e. Unusual shape per 2320.5.4.1).  The braced wall panels
above (2 - 4' long plywood panels) terminate on a 5.125 x 16.5 glu-lam
beam
over the Garage.  The lateral resisting element used below is a single
cantilever steel tube column (attached to the glu-lam).  The steel
column
being considered as a "nonconventional" structural element (it's not
wood,
and it's not a braced wall panel).

What is the so-called "rational" force that this cantilever column
should be
designed for?

1. Determine the greater of the tributary wind load and seismic load to
the
cantilever column.  For seismic, using R = 2.2 for this cantilever steel
column only.  Let's ignore the rigid vs. flexible wood diaphragm issue
for
this discussion, OR

2. Determine the number of braced wall panels, in the 1st story, that
"would" be required to meet Section 2320.11.3 (say 2 - 4' long plywood
panels) ... calculate the total allowable load that those 2 panels could
resist ... then design the cantilever steel column for that load.

Obviously, there may be a huge difference between approach #1 & #2.
Personally I have a problem with approach #2.

I'd like to hear other opinions and share them with the design engineer.
You can send responses direct to me as well.

Regards,
Steven T. Hiner, SE
shiner(--nospam--at)folsom.ca.us

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