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Re: 1997 UBC 2320.2 & 1605.2

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Dennis,

Wouldn't all of the elements in the shear line need to be designed to the
R=2.2?  And isn't the R value essential a number corresponding to the
elements ability to disipate energy, which would explain why a cantilevered
column would be so poor?

Pat Clark


-----Original Message-----
From: Dennis Wish <dennis.wish(--nospam--at)verizon.net>
To: seaint(--nospam--at)seaint.org <seaint(--nospam--at)seaint.org>
Date: Friday, February 08, 2002 1:35 PM
Subject: RE: 1997 UBC 2320.2 & 1605.2


>Steve,
>Assuming that the condition occurs in one location (in one line of
>shear) I would assume the R for the home as 5.5 and adjust the one line
>of shear in the following manner:
>1. At the upper floor where you are using (I assume) proprietary braced
>frames - I would adjust the shear in this line for an R of 4.5 - i.e.;
>take the tributary shear at this line and multiply it by 5.5/4.5 to
>adjust its stiffness.
>2. The uplift and compression generated from the braced panels need to
>be designed into the GLB below. The problem is that the flexibility of
>the GLB may contribute to greater deflection in the braced panels above.
>As I recall, the Seismic Design Manual Volume II had a wood problem
>where a shearwall was placed on top of a wooden beam. You might want to
>refer to this and find out if there were any corrections to the method
>posted after publication.
>3. I would then take the accumulated shear at the second floor diaphragm
>(roof and 2nd floor shear)from the original analysis which used R of 5.5
>and adjust this for an R of 2.2 to design the cantilevered column.
>
>This is how I would design each step from roof down to foundation,
>however, while I would do this to satisfy code (or SEAOC's Blue Book
>opinion on cantilevered columns) I don't really understand their intent
>in penalizing a structure because it was not common practice in the past
>to design plywood shearwalls, or embedded columns to deflection limits.
>
>Just for discussion purposes, it does not make much sense to me to
>adjust the R value for each member design. I understand when making a
>connection of steel to steel that increasing the Rw value from the old
>UBC (94 version) assured a sufficiently larger factor of safety in the
>design of the connection. However, with embedded or cantilevered
>columns, the stiffness issue is moot if you are designing the member for
>deflection rather than allowable bending. What I mean is that while you
>may still be within the allowable stress range for bending, the column
>can still deflect more than the code limit (assuming 0.0025H). This was
>the problem that occurred when engineers designed high load plywood
>shearwalls by assuming that they only needed to comply with a 3.5:1
>aspect ratio (H/b). When these walls were studied after failures, it
>became apparent that the allowable wall deflection occurred at a much
>lower capacity than the wall was sheathed for.
>
>So why not just require the design of the plywood wall, braced frame or
>cantilevered column within an allowable deflection limit based on the
>initial demand calculated to the entire building (say R of 5.5)? What
>benefit does it serve to design the columns to an R of 2.2 when it only
>makes them much stiffer than they need to be?
>
>Does this make much sense? It is understandable to me when designing a
>connection - but not when designing a shear resisting element to a set
>deflection limit.
>
>Dennis S. Wish, PE
>California Professional Engineer
>The Structuralist.Net Information Infrastructure
>
>Website:
>http://www.structuralist.net
>
>."The truly educated never graduate"
>-----Original Message-----
>From: Steve Hiner [mailto:shiner(--nospam--at)folsom.ca.us]
>Sent: Friday, February 08, 2002 12:04 PM
>To: 'seaint(--nospam--at)seaint.org'
>Subject: 1997 UBC 2320.2 & 1605.2
>
>Section 2320.2/Design of Portions - essentially states that
>"nonconventional" structural elements are to be designed in accordance
>with
>Section 1605.2
>
>Section 1605.2 requires a "rational analysis in accordance with well
>established principles of mechanics ... etc."
>
>Here's the problem:  two story wood frame residence with braced wall
>panels
>in the 2nd story that are not continuous through the 1st story to the
>foundation (i.e. Unusual shape per 2320.5.4.1).  The braced wall panels
>above (2 - 4' long plywood panels) terminate on a 5.125 x 16.5 glu-lam
>beam
>over the Garage.  The lateral resisting element used below is a single
>cantilever steel tube column (attached to the glu-lam).  The steel
>column
>being considered as a "nonconventional" structural element (it's not
>wood,
>and it's not a braced wall panel).
>
>What is the so-called "rational" force that this cantilever column
>should be
>designed for?
>
>1. Determine the greater of the tributary wind load and seismic load to
>the
>cantilever column.  For seismic, using R = 2.2 for this cantilever steel
>column only.  Let's ignore the rigid vs. flexible wood diaphragm issue
>for
>this discussion, OR
>
>2. Determine the number of braced wall panels, in the 1st story, that
>"would" be required to meet Section 2320.11.3 (say 2 - 4' long plywood
>panels) ... calculate the total allowable load that those 2 panels could
>resist ... then design the cantilever steel column for that load.
>
>Obviously, there may be a huge difference between approach #1 & #2.
>Personally I have a problem with approach #2.
>
>I'd like to hear other opinions and share them with the design engineer.
>You can send responses direct to me as well.
>
>Regards,
>Steven T. Hiner, SE
>shiner(--nospam--at)folsom.ca.us
>
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