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RE: 1997 UBC 2320.2 & 1605.2

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My understanding of this plan was that the upper garage braced wall line
sits on a beam.  This is an unusual shape if the setback does not meet the
2320.5.4.1 exception. If constructed per one this exception then you do not
need to design a cantilevered column for the original condition described.

If the exterior walls do not stack and it does meetthe 2320.5.4.1 exception,
the building is unusually shaped,and the 2320.2 design of portions does not
apply, and the the whole building  needs to be engineered per 2320.5.4.
2320.5.4 plainly reads "when of unusual shape buildings ... shall have a
lateral-force-resisting system designed to resist forces in Chapter 16".
Doesn't leave much to interpretation.  Also 1630.4.4 requires the use of the
lowest R [cantilevered column R=2.2] in a direction for the entire building
when there is a combination of systems in that direction.  You have one
cantilevered column in an "unusually shaped" building and the whole building
would have to be designed for R=2.2 in that direction.

If this plan is being built in Anchorage, please tell me the permit number
so that I can pull it and require the whole building to be engineered the
way 97 UBC 2320.5.4, and previous and subsequent codes require for irregular
buildings.



-----Original Message-----
From: Pat Clark [mailto:bcinc(--nospam--at)nanosecond.com]
Sent: Saturday, February 09, 2002 7:38 AM
To: seaint(--nospam--at)seaint.org
Subject: Re: 1997 UBC 2320.2 & 1605.2


Dennis,

Wouldn't all of the elements in the shear line need to be designed to the
R=2.2?  And isn't the R value essential a number corresponding to the
elements ability to disipate energy, which would explain why a cantilevered
column would be so poor?

Pat Clark


-----Original Message-----
From: Dennis Wish <dennis.wish(--nospam--at)verizon.net>
To: seaint(--nospam--at)seaint.org <seaint(--nospam--at)seaint.org>
Date: Friday, February 08, 2002 1:35 PM
Subject: RE: 1997 UBC 2320.2 & 1605.2


>Steve,
>Assuming that the condition occurs in one location (in one line of
>shear) I would assume the R for the home as 5.5 and adjust the one line
>of shear in the following manner:
>1. At the upper floor where you are using (I assume) proprietary braced
>frames - I would adjust the shear in this line for an R of 4.5 - i.e.;
>take the tributary shear at this line and multiply it by 5.5/4.5 to
>adjust its stiffness.
>2. The uplift and compression generated from the braced panels need to
>be designed into the GLB below. The problem is that the flexibility of
>the GLB may contribute to greater deflection in the braced panels above.
>As I recall, the Seismic Design Manual Volume II had a wood problem
>where a shearwall was placed on top of a wooden beam. You might want to
>refer to this and find out if there were any corrections to the method
>posted after publication.
>3. I would then take the accumulated shear at the second floor diaphragm
>(roof and 2nd floor shear)from the original analysis which used R of 5.5
>and adjust this for an R of 2.2 to design the cantilevered column.
>
>This is how I would design each step from roof down to foundation,
>however, while I would do this to satisfy code (or SEAOC's Blue Book
>opinion on cantilevered columns) I don't really understand their intent
>in penalizing a structure because it was not common practice in the past
>to design plywood shearwalls, or embedded columns to deflection limits.
>
>Just for discussion purposes, it does not make much sense to me to
>adjust the R value for each member design. I understand when making a
>connection of steel to steel that increasing the Rw value from the old
>UBC (94 version) assured a sufficiently larger factor of safety in the
>design of the connection. However, with embedded or cantilevered
>columns, the stiffness issue is moot if you are designing the member for
>deflection rather than allowable bending. What I mean is that while you
>may still be within the allowable stress range for bending, the column
>can still deflect more than the code limit (assuming 0.0025H). This was
>the problem that occurred when engineers designed high load plywood
>shearwalls by assuming that they only needed to comply with a 3.5:1
>aspect ratio (H/b). When these walls were studied after failures, it
>became apparent that the allowable wall deflection occurred at a much
>lower capacity than the wall was sheathed for.
>
>So why not just require the design of the plywood wall, braced frame or
>cantilevered column within an allowable deflection limit based on the
>initial demand calculated to the entire building (say R of 5.5)? What
>benefit does it serve to design the columns to an R of 2.2 when it only
>makes them much stiffer than they need to be?
>
>Does this make much sense? It is understandable to me when designing a
>connection - but not when designing a shear resisting element to a set
>deflection limit.
>
>Dennis S. Wish, PE
>California Professional Engineer
>The Structuralist.Net Information Infrastructure
>
>Website:
>http://www.structuralist.net
>
>."The truly educated never graduate"
>-----Original Message-----
>From: Steve Hiner [mailto:shiner(--nospam--at)folsom.ca.us]
>Sent: Friday, February 08, 2002 12:04 PM
>To: 'seaint(--nospam--at)seaint.org'
>Subject: 1997 UBC 2320.2 & 1605.2
>
>Section 2320.2/Design of Portions - essentially states that
>"nonconventional" structural elements are to be designed in accordance
>with
>Section 1605.2
>
>Section 1605.2 requires a "rational analysis in accordance with well
>established principles of mechanics ... etc."
>
>Here's the problem:  two story wood frame residence with braced wall
>panels
>in the 2nd story that are not continuous through the 1st story to the
>foundation (i.e. Unusual shape per 2320.5.4.1).  The braced wall panels
>above (2 - 4' long plywood panels) terminate on a 5.125 x 16.5 glu-lam
>beam
>over the Garage.  The lateral resisting element used below is a single
>cantilever steel tube column (attached to the glu-lam).  The steel
>column
>being considered as a "nonconventional" structural element (it's not
>wood,
>and it's not a braced wall panel).
>
>What is the so-called "rational" force that this cantilever column
>should be
>designed for?
>
>1. Determine the greater of the tributary wind load and seismic load to
>the
>cantilever column.  For seismic, using R = 2.2 for this cantilever steel
>column only.  Let's ignore the rigid vs. flexible wood diaphragm issue
>for
>this discussion, OR
>
>2. Determine the number of braced wall panels, in the 1st story, that
>"would" be required to meet Section 2320.11.3 (say 2 - 4' long plywood
>panels) ... calculate the total allowable load that those 2 panels could
>resist ... then design the cantilever steel column for that load.
>
>Obviously, there may be a huge difference between approach #1 & #2.
>Personally I have a problem with approach #2.
>
>I'd like to hear other opinions and share them with the design engineer.
>You can send responses direct to me as well.
>
>Regards,
>Steven T. Hiner, SE
>shiner(--nospam--at)folsom.ca.us
>
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