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RE: 1997 UBC 2320.2 & 1605.2

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Pat, The R value of 2.2 would be designed for the full line of shear -
but only at the level that the embedded cantilever columns occur. I
don't see the purpose of reducing R at the second level that contains
either a conventional Shear Wall or a Braced Frame (I think I suggest
altering R for 4.5 which is appropriate for a braced frame). However, I
also suggested that you alter the shear in the first floor shear
elements in that line of shear for an R of 2.2 which would be cumulative
- so you would be increasing the shear of combined demand from the roof
and second floor diaphragm.
It's been a lot years since I have been in school. When you speak of
energy dissipation I assume you are speaking of how many cycles it takes
a shear element to reach equilibrium. Does it really matter as long as
the element is not allowed to continue to cycle. The goal is not to
exceed the allowable deflection and in my opinion, stiffness is related
to deflection.

Remember that the concern of cantilevered column originated from an
invalid assumption that the front of the Northridge Meadows Apartments
was weak because it was supported by cantilevered or fixed base columns.
However, sometime later it was reported by Bob Kazanji (sorry about the
spelling Bob) at UC Irvine that the columns at the front of the garage
entry were not fixed but pinned columns. This means that the defect was
a soft-story or open front. I would then assume that the original
engineer designed the lower level cripple walls between the top of
foundation at grade to the first floor diaphragm on three sides would
resolve the shear caused by rotation due to the lack of shear in the
open-front. This proved many years before to bee inadequate (my first
memory of this was after the Whittier Narrows Earthquake when the City
of Los Angeles rejected rotational design whenever a living structure
occurred above an open-front or soft-story design. (the difference in my
use of terms is that an open-front has no shear or is cantilevered some
distance from the nearest parallel shearwall while a soft-story provide
shear resistance, but the panel is too flexible and the deflection
exceeds code allowable story drift.)

I can't answer your question as it relates to the dissipation of energy,
but would think that if this is not a proprietary shear-resisting
element, the time to dissipate energy is only an estimate. IMO our focus
should be on designing to deflection and we need not worry about
increasing shear to the element by reducing R (I stated it incorrectly
in my original post) which only penalizes the client by inflating his
cost of materials and limiting deflection to much less than the designed
value by increasing the member stiffness required to resist an increased
demand.

This is far from an exact science here.

Dennis S. Wish, PE
California Professional Engineer
The Structuralist.Net Information Infrastructure

."The truly educated never graduate"
-----Original Message-----
From: Pat Clark [mailto:bcinc(--nospam--at)nanosecond.com]
Sent: Saturday, February 09, 2002 8:38 AM
To: seaint(--nospam--at)seaint.org
Subject: Re: 1997 UBC 2320.2 & 1605.2

Dennis,

Wouldn't all of the elements in the shear line need to be designed to
the
R=2.2?  And isn't the R value essential a number corresponding to the
elements ability to disipate energy, which would explain why a
cantilevered
column would be so poor?

Pat Clark


-----Original Message-----
From: Dennis Wish <dennis.wish(--nospam--at)verizon.net>
To: seaint(--nospam--at)seaint.org <seaint(--nospam--at)seaint.org>
Date: Friday, February 08, 2002 1:35 PM
Subject: RE: 1997 UBC 2320.2 & 1605.2


>Steve,
>Assuming that the condition occurs in one location (in one line of
>shear) I would assume the R for the home as 5.5 and adjust the one line
>of shear in the following manner:
>1. At the upper floor where you are using (I assume) proprietary braced
>frames - I would adjust the shear in this line for an R of 4.5 - i.e.;
>take the tributary shear at this line and multiply it by 5.5/4.5 to
>adjust its stiffness.
>2. The uplift and compression generated from the braced panels need to
>be designed into the GLB below. The problem is that the flexibility of
>the GLB may contribute to greater deflection in the braced panels
above.
>As I recall, the Seismic Design Manual Volume II had a wood problem
>where a shearwall was placed on top of a wooden beam. You might want to
>refer to this and find out if there were any corrections to the method
>posted after publication.
>3. I would then take the accumulated shear at the second floor
diaphragm
>(roof and 2nd floor shear)from the original analysis which used R of
5.5
>and adjust this for an R of 2.2 to design the cantilevered column.
>
>This is how I would design each step from roof down to foundation,
>however, while I would do this to satisfy code (or SEAOC's Blue Book
>opinion on cantilevered columns) I don't really understand their intent
>in penalizing a structure because it was not common practice in the
past
>to design plywood shearwalls, or embedded columns to deflection limits.
>
>Just for discussion purposes, it does not make much sense to me to
>adjust the R value for each member design. I understand when making a
>connection of steel to steel that increasing the Rw value from the old
>UBC (94 version) assured a sufficiently larger factor of safety in the
>design of the connection. However, with embedded or cantilevered
>columns, the stiffness issue is moot if you are designing the member
for
>deflection rather than allowable bending. What I mean is that while you
>may still be within the allowable stress range for bending, the column
>can still deflect more than the code limit (assuming 0.0025H). This was
>the problem that occurred when engineers designed high load plywood
>shearwalls by assuming that they only needed to comply with a 3.5:1
>aspect ratio (H/b). When these walls were studied after failures, it
>became apparent that the allowable wall deflection occurred at a much
>lower capacity than the wall was sheathed for.
>
>So why not just require the design of the plywood wall, braced frame or
>cantilevered column within an allowable deflection limit based on the
>initial demand calculated to the entire building (say R of 5.5)? What
>benefit does it serve to design the columns to an R of 2.2 when it only
>makes them much stiffer than they need to be?
>
>Does this make much sense? It is understandable to me when designing a
>connection - but not when designing a shear resisting element to a set
>deflection limit.
>
>Dennis S. Wish, PE
>California Professional Engineer
>The Structuralist.Net Information Infrastructure
>
>Website:
>http://www.structuralist.net
>
>."The truly educated never graduate"
>-----Original Message-----
>From: Steve Hiner [mailto:shiner(--nospam--at)folsom.ca.us]
>Sent: Friday, February 08, 2002 12:04 PM
>To: 'seaint(--nospam--at)seaint.org'
>Subject: 1997 UBC 2320.2 & 1605.2
>
>Section 2320.2/Design of Portions - essentially states that
>"nonconventional" structural elements are to be designed in accordance
>with
>Section 1605.2
>
>Section 1605.2 requires a "rational analysis in accordance with well
>established principles of mechanics ... etc."
>
>Here's the problem:  two story wood frame residence with braced wall
>panels
>in the 2nd story that are not continuous through the 1st story to the
>foundation (i.e. Unusual shape per 2320.5.4.1).  The braced wall panels
>above (2 - 4' long plywood panels) terminate on a 5.125 x 16.5 glu-lam
>beam
>over the Garage.  The lateral resisting element used below is a single
>cantilever steel tube column (attached to the glu-lam).  The steel
>column
>being considered as a "nonconventional" structural element (it's not
>wood,
>and it's not a braced wall panel).
>
>What is the so-called "rational" force that this cantilever column
>should be
>designed for?
>
>1. Determine the greater of the tributary wind load and seismic load to
>the
>cantilever column.  For seismic, using R = 2.2 for this cantilever
steel
>column only.  Let's ignore the rigid vs. flexible wood diaphragm issue
>for
>this discussion, OR
>
>2. Determine the number of braced wall panels, in the 1st story, that
>"would" be required to meet Section 2320.11.3 (say 2 - 4' long plywood
>panels) ... calculate the total allowable load that those 2 panels
could
>resist ... then design the cantilever steel column for that load.
>
>Obviously, there may be a huge difference between approach #1 & #2.
>Personally I have a problem with approach #2.
>
>I'd like to hear other opinions and share them with the design
engineer.
>You can send responses direct to me as well.
>
>Regards,
>Steven T. Hiner, SE
>shiner(--nospam--at)folsom.ca.us
>
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