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RE: Rigid vs. Flexible Diaphragm

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Let me take this step by step so there is no misunderstanding:

1. My original problem and my design intention was not for high-wind
regions. The designs in my area are one-story (under 18-feet to peak of roof
as this is a resort area and the view is maintained under local zoning for
single family residences in most areas of the Coachella Valley) structures
with less than 90-mph (not including) and exposure C. 
2. Conventional construction, although not relevant to an engineered design,
but maintained due to the low event or non-event of failure, only requires
double trimmers and double king studs when openings exceed 8-feet. There is
no limit to the amount or width of openings as long as the provisions for
braced panel (no Holddowns required if aspect ratio's are maintained)
spacing are compliant.
3. I did not suggest that you shouldn't consider the hinge, but let's look
at my example - 16'-0" wide header (8'-0" opening + 7'-0" for the Hardy
Frame width) 12'-0" to bottom of header and 14'-0" to double top plate
(finished slab to finished slab). The wind pressure for this area under
15'-0" is 70 mph and Exposure C, the reaction at the end of the header
considering a 7'-0" tributary wind normal to the header is 974-lb located
12'-0" from the slab and 2'-0" from the double top plate. 
Using 2-2x6 DF#1 KD Studs (or a 3x6 or 4x6) the king stud will be 11% below
maximum allowable stress (we tend to use #1 lumber for posts and in some
cases for all studs if it is easier to find here - especially in Kiln Dried
lumber since we are in the desert.

This does point to the fact that as the width of the header increases and
the wind load exceeds 90-mph or exposure C OR if the larger width of the
opening occurs at higher elevations on a multi-story building that you
should check out the forces against the king post.

3. You made a comment:

"FYI, I try to stay away from Hardy frames because they are very stiff
compared to plywood walls and strong walls.  When using RDA, these frames
attract too much load."

You hit the nail on the head here - they do attract load. This is the reason
for using them as they fit into spaces where conventional plywood walls do
not work. I would not arbitrarily use Hardy Frame, ShearMax or Strongwalls
where I could construct a traditional plywood shearwall - it is not
economical.  Furthermore, you are not likely to use only RDA for
light-framing alone. If you do, you leave yourself liable for potential
open-front or soft-story failures. The code (at least the 97 UBC) does not
say you can't use an open-front design - it's just that historically we know
in Southern California that it does not work - especially where there is a
living structure above. If you attempt to balance based on the worst case
results from Flexible (wind and seismic) and Rigid then you need to reverse
engineer the RDA to push the shear back into the diaphragm for it to be
distributed to other walls. This can cause a domino effect - especially if,
as you pointed out, you introduce a stiffer wall such as a Hardy frame into
the model.)

4. You said; "I do not see how the partitions would support the out-of-plane
forces applied on the Hardy frame.”

Let me see if I can explain this: What matters is the "UNSUPPORTED" length
of the hinge. Don't forget we haven't considered discussing out-of-plane
bending on long headers since these are unsupported at all heights between
trimmer posts. When was the last time you calculated a 16'-0" header in
forces normal to the all? I don't think I ever had and maybe I should start!
However, wherever sheathing is used near or at the ends of long headers, I
strap vertically and horizontally to blocking on each side of the header to
stiffen the corners and reduce the typical shear cracking that is notable in
older homes without these connections

When the opening is closed (closed window, door, sliding door etc) the wind
is acting tributary to the area of the header. Introducing a Hardy panel -
even if connected only to the header - is still introducing a solid wall
with a hinge from foundation to roof (or upper plate) - it never opens and
is sheathed over with some finish or another. 

Now if you place an interior wall flush to the inside and perpendicular to
the Hardy Panel continuous from floor to diaphragm above, the interior wall
acts to brace the panel and the header. You might point out that there is no
positive connection of interior non-bearing partitions to the diaphragm
above, but I'll argue that there is by nature of the connection of the
finishing materials - even if you assume 25-plf for gypsum, there will
generally be enough resistance to brace the exterior wall out-of-plane.

It is more critical where the opening is large (like a garage door) and no
panels are introduced. By the way, look at the Hardy details they provide
and which have been introduced into their COLA and ICBO report that
demonstrate a connection of the Hardy Panels at each side of a garage with
the garage header extended over the top of the panels.

5. In you next statement I think you might have misunderstood something so I
apologize if I am wrong - please correct me:
"IMO, both are important.   If you use a shorter hardy frame that does not
extend to the diaphragm, you have to reduce the shear capacity to account
for the overturning forces that is transferred from the cripple wall above
the frame.  Also, you have to account for the additional deflection due to
higher overturning forces.  You also need to provide means of transferring
the overturning force from the cripple wall to the hardy frame."

Let me start simply - I would never sheath only the width of the shearwall
or Hardy Frame from plate to plate. When I said I would sheath the wall
above the header I meant from one end of the building to the other - the
length of the lap spliced double top plate - no less. It is important to
make sure that the horizontal edge of the plywood is blocked adjacent to the
ends of the headers. If the edge nailing of the plywood at the header can
not sufficiently meet the drag capacity, the strapping over the blocked area
should be sufficient to develop more capacity than is needed.

If you do this, there rocking of the Hardy Panel is of more concern than any
possible uplift or rocking of the pony wall. This is virtually the same
argument that is used when connecting a Gable end truss in shear. Some
suggest sheathing the truss with plywood, others have the truss designed for
the reaction at the end wall and still others find that per linear foot, the
metal lath and stucco is sufficient to develop more than enough capacity
even at 90-plf. In-plane shear is rarely that high when you consider the
full depth of the diaphragm and the boundary nailing unless the aspect ratio
of the horizontal diaphragm encroaches upon 4:1 (long and narrow).

6. " In order to transfer the load to adjacent members, the wall skin
(plywood, plaster) must have adequate stiffness to be able to transfer the
load.  Plywood (plaster, drywall, etc) are not stiff in the out-of-plane

I would agree with you on this. My point prior to this was that for normal
wind loads (70 mph and Exposure C) and traditional height walls (8 to 12
foot plate to plate), a double king stud, a 3x, 4x or 6x post will be
sufficient and manufactured clips will secure the post from plate to plate.

However, what we don't consider is the "system". There has been no attempt
that I know of to define and test the wrapping as a system acting
out-of-plane. Personally, I think that the lack of damaged to tall vaulted
ceiling engineered homes in past earthquakes is indicative of the fact that
we have not addressed the building as a system but rather as components. A
couple of examples might be:
a) Slab on grade foundation and the effect that the 3-1/2" slab has to
resist uplift on shearwalls connected to the thickened slab edge,
b) The effect of strapping over headers and plywood sheathing on a wall to
resist this "hinge effect" we talked about. 
c) How interior partitions dampen the diaphragm movement as it does in URM
design (special procedure). 
d) Floor diaphragm deflection on post and pier foundations 18" above grade.

I suspect I'll be retired before many of these questions have been answered
but I hope that what I replied to above clears up anything you might have
misinterpreted from my original posts. All in all, I don't think we are too
far apart in our thinking about the hinge effects. The majority of wood
design is, under the 97 UBC, done with ASD design and there is a sufficient
factor of safety that I think we ignore (and possibly rightly so) that gives
us some padding in our professional judgment. For example, I am retrofitting
a 1920 concrete fire station that is 24 feet wide and the roof was converted
to a floor load in the past 50 years and used to house the firemen for
sleeping and living day and night. The floor joists are full 2"x10.5" deep
dry joists at 16-inch on center and based on the floor live load of even
40-psf, the joists are considered 30+% overstressed using pre-1991 Wood
design stresses. When looking at the exposed framing, there is no deflection
noticeable, no failures when rightly there should be and luck has been with
these men for many years. Put a few too many people on a porch in Chicago
and it comes down. Go figure! (Pun intended)

When dealing with monolithic materials we find it easier to calculate the
activity of the material in-plane, out-of-plane and from gravity load.
Introduce steel into concrete and it becomes a bit more difficult, but
introduces all of the different materials and proprietary ones at that on
wood and the matrix become infinitely more complicated. I think we are babes
in this area and need a lot of years to mature in our understanding of
issues as simple as this hinge effect.

Go for it :>)


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